Seismic response of a motorway bridge founded in an active landslide: a case study

Abstract

A twin girder 7-pier bridge, belonging to the "Egnatia" highway that has been facing numerous challenging geohazards, is built within an active landslide. Its seismic performance is investigated here through a comprehensive analysis of the interaction between bridge, foundation, and the precarious slope, which might affect 4 of the piers. The numerical 3D modeling considers in a realistic way the coupled effects of topography, soil nonlinearity, slope instability, and reinforced-concrete plasticity during seismic loading (kinematic and inertial). Alternative foundation schemes and slope stabilizing techniques are generically compared and evaluated. The aim is to develop a multi-hazard risk assessment platform that could facilitate the long-term management of motorways while shedding some light on the multi-hazard soil-structure interaction (MH-SSI).

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Bulletin of Earthquake Engineering
https://doi.org/10.1007/s10518-022-01544-3
1 3
ORIGINAL ARTICLE
Seismic response ofamotorway bridge founded inanactive
landslide: acase study
AntoniosMantakas1 · AngelosTsatsis2· MariannaLoli2· RallisKourkoulis1,2·
GeorgeGazetas1
Received: 4 November 2021 / Accepted: 16 October 2022
© The Author(s) 2022
Abstract
A twin girder 7-pier bridge, belonging to the "Egnatia" highway that has been facing
numerous challenging geohazards, is built within an active landslide. Its seismic perfor-
mance is investigated here through a comprehensive analysis of the interaction between
bridge, foundation, and the precarious slope, which might affect 4 of the piers. The numeri-
cal 3D modeling considers in a realistic way the coupled effects of topography, soil non-
linearity, slope instability, and reinforced-concrete plasticity during seismic loading (kin-
ematic and inertial). Alternative foundation schemes and slope stabilizing techniques are
generically compared and evaluated. The aim is to develop a multi-hazard risk assessment
platform that could facilitate the long-term management of motorways while shedding
some light on the multi-hazard soil-structure interaction (MH-SSI).
Keywords Landslides· Multi-hazard· Soil–structure interaction· Nailing piles· Caisson
foundation
* Antonios Mantakas
mantakasantonis@gmail.com; antonismantakas@mail.ntua.gr
Angelos Tsatsis
tsatsis.angelos@gmail.com
Marianna Loli
m.loli@grid-engineers.com
Rallis Kourkoulis
ralisko@yahoo.gr
George Gazetas
gazetas@mail.ntua.gr
1 Department ofCivil Engineering, National Technical University ofAthens, 9, Iroon Polytechniou
Str, Athens, Greece
2 Grid Engineers, Pampouki 3, N., 15451Psychiko, Greece
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1 Introduction
Transportation networks constitute a vital ingredient of modern, densely populated socie-
ties, where the transfer of goods and people bring social and economic prosperity. Motor-
ways are spatially distributed systems comprising bridges, tunnels, culverts, and earth-
works that might be affected by multiple hazards. The smooth operation of these network
systems after a disruptive event is crucial. Bridges are perhaps the most vulnerable assets
of a road network; their damage has been historically considered a significant threat to
motorists (Priestley et al. 1996). Earthquakes and geohazards such as landslides, debris
flow, and ground movement could be a detrimental combination for highway bridges (Bhat-
tacharya etal. 2018; Kawashima and Buckle 2013; Winter etal. 2013). Bridges often cross
streams and slopes, and could be susceptible to erosion, landslides, and liquefaction, all
contributing to damage, with direct and indirect losses (Kawashima etal. 2011). Therefore,
it is important to assess the effects of topography, local soil conditions, potential ground
instabilities, and seismic shaking. This calls for confronting seismic and geotechnical haz-
ards simultaneously (Kiremidjian etal. 2007).
In the last years, the focus on natural hazards and their effects on infrastructure grew
significantly, leading to seismic risk assessment studies of road networks and bridges.
Probabilistic seismic analyses have been adopted to identify the hazard levels, evaluate the
vulnerability of each structural component to each specific hazard, conduct a risk assess-
ment of the assets, and finally assess the resilience of highway networks (Argyroudis etal.
2015; Kilanitis and Sextos 2019; Kiremidjian etal. 2007). But most of these studies have
addressed only the dominant hazard, without considering multi-hazard scenarios that may
affect the infrastructure. The seismic hazard in particular, with emphasis on soil-founda-
tion-superstructure interaction, has been very extensively investigated (I. Anastasopoulos
etal. 2015; Kappos and Sextos 2001; Makris etal. 1994, 1996; Mylonakis et al. 1997).
Hence, the need for a comprehensive approach that considers the simultaneous occurrence
of multiple-hazard actions (Chulahwat and Mahmoud 2017; Gidaris etal. 2017; Yanweera-
sak etal. 2018). More specifically, the simultaneous seismic shaking and ground-failure
actions have been investigated by (Fotopoulou and Pitilakis 2017; Pitilakis and Kalliopi
2013; Tang etal. 2020). In this paper, a comprehensive numerical investigation of the seis-
mic behaviour of a bridge founded on an active landslide has been undertaken to provide
insight on the effects of coupling of seismic and ground kinematics.
Among the methodologies developed the to study the multiple-hazard seismic
soil–foundation one could mention analytical solutions (Di Laora etal. 2017; Elahi etal.
2010; Gazetas etal. 1993; Wen etal. 2015), numerical studies (Barla 2018; Jin etal. 2010;
Kourkoulis etal. 2010; Luo etal. 2019; Uzuoka et al. 2007), and 1-g seismic and cen-
trifuge experiments (Wang and Zhang 2014; Yan etal. 2020). To mitigate landslide risk
and protect the affected infrastructure, the use of stabilising piles and deep foundations has
been examined by (He etal. 2015; Kanagasabai etal. 2011; Kourkoulis etal. 2011, 2012;
Yu etal. 2015). Most of this research investigated soil–structure interaction by decoupling
the superstructure from the foundation in a multi-step approach.
In this study we examine a bridge of the Egnatia motorway in Greece, which is sup-
ported on pile and caisson foundations on top of an active landslide. A detailed numeri-
cal model of the whole structure-soil-foundation system is generated. Rational constitutive
models are adopted to represent both soil and concrete under static and dynamic actions.
A multi-hazard approach is considered herein, where seismic and ground-instability are
imposed simultaneously on the system. The barely acceptable performance obtained from
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the analyses of the foundation–bridge system prompted the investigation of certain slope
stabilisation interventions, in the form of a double series of staggered rows of rigidly-
capped piles. It is shown that with such improvements even a smaller size of foundation
would have offered an acceptable performance, even with the strongest seismic excita-
tion—a refutation of the fallacy commonly held that by conservatively oversizing foun-
dations we achieve greater safety (Ioannis Anastasopoulos etal. 2010). Of course, other
mitigation schemes could also be incorporated for a technically sound solution, avoiding
the unnecessary conservatism in foundation design.
2 The case oftheG1 bridge oftheEgnatia Motorway, Greece
2.1 Summary ofdesign
This bridge (“Panagia” Interchange) of the Egnatia motorway in Greece is a typical
T-girder beam bridge built in 2007. It consists of two branches with 37.5m span and pier
height up to 25m. It is a modern structure with decks supported on elasto-metallic rubber
bearings providing seismic isolation. Part of the bridge is founded on a precarious slope
suffering from creeping deformations, and vulnerable to permanent sliding in a strong seis-
mic event, as well as after heavy rainfall (Fig.1).
Seven circular piers support each deck and are founded on deep foundations. A longitu-
dinal section of the bridge is given in Fig.2. The foundations are either pile groups of 6–8
piles of 1.2m diameter and 23–30m length (P1–P4), or massive caissons (P5–P7). The
latter are located within the area of potential landslide, not only to ensure sufficient founda-
tion capacity and protect the structure, but also to stabilize the slope. The depth of these
caissons varies between 20 and 32m below ground level and their diameter between 5 and
7m. Twenty additional 0.8m–diameter piles were constructed around the caisson periph-
ery perhaps for enhancing the structural capacity of the foundation (or other unknown to
Fig. 1 Photo of G1 Bridge of Egnatia motorway
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us reasons related to construction). Typical foundation cross-sections are shown in Fig.3,
while the geometric parameters are given in Table1.
2.2 Geotechnical conditions
Figure4 presents a lateral cross-section of the slope showing the typical geotechnical pro-
file with the twin piers P5 and their caisson foundation. The parameters and geotechni-
cal characteristics of the soil layers are summarized in Table2, based on the geotechnical
investigation by "Geognosi", a geotechnical engineering company.
Fig. 2 Longitudinal section of the left branch of bridge G1
Fig. 3 Typical cross-section of the P1–P4 pile groups and of the P5–P7 caisson foundations
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Table 1 Foundation geometry
parameters Parameters P1 P2 P3 P4 P5 P6 P7
h (m) 11 16 23 25 20 20 17
Bt (m) 8 8 10 12 – – –
Lt (m) 10 10 10 12 – – –
No. of piles 6 6 8 8 20 18 14
Lp (m) 30 26 23 23 – – –
dp (m) 32 27 20
d1 (m) 11 11 6
Dt (m) – – – – 9 8 7
Dc (m) – – – – 7 6 5
Dp (m) 1.2 1.2 1.2 1.2 0.8 0.8 0.8
Fig. 4 Typical geotechnical/geological cross-section of the bridge and slope showing the groundwater table
(blue dotted line), the sliding surface (yellow line), and the formations encountered on site
Table 2 Summary of geotechnical soil properties of the soil units encountered within the study area
Soil Unit VIII IV
SPT count number NSPT 16 (14–19) 43 –
Shear-wave velocity (from SPT
calculation)
Vs [m/s2] 244 339 > 800
1-D Compression modulus Ds [MPa] – 50 > 900
Water content w (%) 1.1 11.8 –
Unit weight γ [kN/m3] 21 24 25
Void ratio e0.439 – 0.439
Effective friction angle φ’ [deg] 33 24 35
Effective cohesion c’ [kPa] 5 35 70
Residual friction angle φ’r18 20 –
Undrained shear strength Su [kPa] 70 220 –
Coefficient of Permeability k [cm/s] 4.6 × 10−5 2.4 × 10−5 1 × 10−5
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3 The multi‑hazard environment
3.1 Seismicity
The G1 bridge is located in a mountainous region of moderate seismicity (design PGA for
475-year return period according to EC8 is 0.16g). In this study, the OpenQuake open-
source software is used for performing the seismic hazard and deaggregation analyses, by
employing the area source model of the SHARE project (Giardini etal. 2013) (Fig.5a) and
the ground motion prediction equation of Boore and Atkinson (2008). The methodology
is described in more detail in Loli et al. (2020). The Intensity Measure (IM) considered
herein is the average peak spectral acceleration, AvgSa (Vamvatsikos and Cornell 2005),
and is calculated using the following equation:
(1)
IM
=AvgSa

TRi

=

n

i=1
Sa

TRi
1∕n
Fig. 5 Seismic hazard assessment and CS-based record selection (Loli et al. 2020): a “SHARE” area
source model for Greece and Egnatia’s segment (green line); b site-specific seismic hazard curve; c CMS,
CMS ± 2σ and response spectra of the 11 selected records
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We notice that IM is derived by combining n spectral acceleration ordinates (Sα) at peri-
ods TRi. Each ordinate is the geometric mean of the 5%-damped spectral acceleration from
the two horizontal components. The period range of [0.3 s, 3.0 s] with an increment of
0.1s is adopted here. The resulting site-specific seismic hazard curve is depicted in Fig.5b
in terms of AvgSa versus the mean annual frequency (MAF) of exceedance. The conditional
mean spectrum (CMS), its 2.5th and 97.5th percentiles (CMS ± 2σ) and the response spec-
tra of the selected records are presented in Fig.5d.
Eleven (11) acceleration time histories for each seismic intensity level (i.e., weak, mod-
erate, and strong) have been adopted as excitation at the base of the models. The accelero-
grams and the elastic response spectra of the “strong” scenario only are presented in Fig.6.
A summary of the characteristics of the selected ground records is provided in Table3.
For this study, a scenario-based approach has been adopted. Seismic motions consistent
with three different levels of the seismic hazard (i.e., weak: probability of exceedance 20%
in 50years; moderate: probability of exceedance 10% in 50years; and strong: probability
of exceedance 2% in 50years) were considered with the current condition of the creeping/
failing slope (as described in detail below). Analysis has been performed for the worst-
case scenario of maximum groundwater height reaching the ground surface—a reasonably
conservative approximation in view of past observations that found the ground water level
above the failure zone even during dry periods, the relatively high precipitation levels in
the area, and the low soil permeability.
In this case, the effect of pore pressures in reducing the effective stress and thereby the
strength of the weak zone is indirectly taken into by proportionally reducing the effective
friction angle (φ’) in the calculations. Results are presented for all the investigated haz-
ard scenarios, but more details and discussion are provided for the most detrimental case:
intense earthquake (probability of exceedance 2% in 50years) and water table is at the
ground surface.
For the most intense level of shaking, results from analysis imply mobilisation of the
failure surface and significant slope movement. Residual displacements of the slope NIN2,
within the failure wedge, range from 0.5m to over 3m depending on the details of the
excitation.
3.2 Potential landslide actions
Monitoring the slope movements around the site began in 2002. Several inclinometers and
a few piezometers were installed at different locations in the slope (Fig.7a). Unfortunately,
most of the instruments installed in 2002 stopped giving readings when construction of
the bridge started in 2007. Additional instruments were installed in boreholes N1–N6 in
2007 and on the bridge IM4–IM6 in 2014. Horizontal displacements recorded by the latter
instruments are plotted in Fig.7b. Accumulated displacements reaching up to 8cm were
measured in location N2, 46m, uphill from Pier 5. Similar displacements were measured
downhill, especially of N5 and N6. But, N6 stopped working in 2007 (shortly after instal-
lation), signifying an abrupt increase of shear deformations (revealed by the high displace-
ment rate). Analysis of inclinometer readings led to the identification of a landslide area
that comprises the central piers P4, P5, P6. Due to its steeply sloping terrain, the cross-
section near P5 pictured in Fig.7a is recognized as the most critical. The unstable soil layer
is between about 8 and 11m thick, and the distributions of horizontal displacements with
depth, as per Fig.7b, indicate strain accumulation in a thin zone (slip plane).
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The stability of piers P5(L) and P5(R) depends on their massive foundations, which were
intended to also act as a slope stabilizing measure. The foundations consist of a 32m deep
reinforced concrete circular caisson diameter 7m in diameter. Their capacity is enhanced by
20 Φ80 cm piles installed in the periphery at a distance from the caisson, with a common rigid
-0.8
-0.4
0
0.4
0.8
010203040
CPM000
-0.8
-0.4
0
0.4
0.8
010203
04
0
CHY101-E
-0.8
-0.4
0
0.4
0.8
010203
04
0
H-CMP015
-0.8
-0.4
0
0.4
0.8
010203040
H-COH000
-0.8
-0.4
0
0.4
0.8
010203040
H-E06140
-0.8
-0.4
0
0.4
0.8
010203040
CLD195
-0.8
-0.4
0
0.4
0.8
010203
04
0
GYN000
-0.8
-0.4
0
0.4
0.8
010203
04
0
SSE330
-0.8
-0.4
0
0.4
0.8
010203040
PSA090
-0.8
-0.4
0
0.4
0.8
010203
04
0
B-CAL225
-0.8
-0.4
0
0.4
0.8
010203
04
0
B-ICC000
0
0.5
1
1.5
2
00.5 11.5
a : ga : ga : ga : ga : g
a : g
Sa : g
t : s
t : s
T : s
Fig. 6 Accelerograms and elastic (ξ = 5%) response spectra of the ansemble of 11 excitations used as input
in the numerical analyses, that are consistent with the seismic hazard with probability of exceedance 2% in
50years (Loli etal. 2020)
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cap. Critical for the design was the seismic excitation. Note that the aforementioned recorded
displacements were solely creeping deformations along the "failure" zone (slip-plane), prob-
ably a result of seasonal movement of the water table (Loli et al. 2020). In addition, recent
information indicated that the abrupt increase of the rate of slope deformation during 2013
(Figs.7b and 12) may be attributed to the fact that excavation material and debris were placed
on top of the slope, increasing detrimentally the weight of the moving mass. Naturally, even
larger displacements might be expected in case of an earthquake.
Table 3 Characteristics of the selected ground records
No. Earthquake Year Station MwPGA (g)
*unscaled Scaling Factor Vs30 (m/s)
1 Cape mendocino 1992 Cape Mendocino/CPM000 7.01 1.50 0.57 567
2 Chi-Chi 1999 CHY101-E 7.62 0.18 1.39 804
3 Coalinga 1983 Parkfield/COH000 6.36 0.14 1.39 384
4 Imperial valley 1979 Compuertas/CMP015 6.53 0.19 3.51 260
5 Imperial valley 1979 El Centro Array6/E06140 6.53 0.41 0.43 210
6 Kocaeli 1999 Goynuk/GYN000 7.51 0.13 2.68 362
7 Loma prieta 1989 CoyoteLakeDam/CLD195 6.93 0.16 1.73 561
8 Northridge 1994 TerminalIsland/SSE330 6.69 0.19 3.31 260
9 N. Palm springs 1986 PalmSpringsAirport/PSA090 6.06 0.19 2.32 312
10 Superstition hills 1987 Cal/B-CAL225 6.54 0.18 1.85 348
11 Superstition hills 1987 El Centro Imp/B-ICC000 6.54 0.36 0.75 192
(a)
G1 Bridge
Rural Road
Egnaa Odos -
Access Road
Landslide Zone
0
10
20
30
40
50
60
70
80
10/06
07/09
04/12
12/14
09/17
06/20
Horizontal Displacement (mm)
G10 NIN4 NIN5 NIN6
NIN2 NIN3 NIN1
(b)
Fig. 7 Illustration of a topographic map indicating the location of inclinometers, landslide zone, and distri-
bution of horizontal displacements (NIN2 readings), b results of all installed inclinometer and evolution of
displacements with time
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4 Numerical modelling oflandslide–foundation interaction
underseismic excitation
4.1 Initial “As‑Built” model
An analysis framework was developed to obtain the slope movement and pier performance
of the bridge. The finite element code ABAQUS is employed in a series of non-linear seis-
mic analyses. Pier 5L is the most vulnerable to the combined slope and seismic loading,
due to its location within the landslide zone (Fig.2). The axis-to-axis distance of the cais-
sons of the two adjacent piers (belonging to the twin branches of the bridge) is about 3.5
diameters. For dynamic inertial loading at their top under elastic conditions in homogene-
ous soil the interaction between the two caissons would be significant (Dobry and Gaze-
tas 1988). However, soil nonlinearity that develops in this case, drastically reduces such
interaction (Gazetas etal. 2021; Kanellopoulos and Gazetas 2020; Radhima etal. 2021).
Moreover, the “kinematic” effects are prominent in our case, because in addition to the
direct seismic wave loading there is indirect loading from the lateral soil movement. As
is well known, interaction between piles due to kinematic loading is much less significant
even in elastic soil (Gazetas and Mylonakis 1998; Kaynia and Kausel 1982). Therefore,
we do not expect any substantial interaction effects between the two adjacent caissons of
the two branches of the bridge, especially in view of intense soil nonlinearity. The three-
dimensional FE modelling realistically captures the soil–foundation interaction, accounting
for the exact geometry of soil and superstructure, kinematic boundaries, and pile group-
caisson response.
Using a detailed 3D model simulating the caisson-pile P5 foundation we look for the
damage due to a potential slope instability originating during strong earthquake shaking
(Figs.8 and 9). Inelastic behaviour of soil, foundation, and column are modelled explic-
itly while the deck remains elastic. The model encompasses 640m of ground in the lon-
gitudinal direction (of seismic shaking) and 45 m deep ground below the caisson toe.
The adopted model comprises the full geometry of the slope/valley, and displacement is
allowed only in the longitudinal direction of y axis. Half the model is shown in Fig.8 for
clarity. Dynamic analyses are performed for all seismic scenarios. The complete foun-
dation-soil-superstructure is modelled for the most critical cross-section. In the analysis,
acceleration time histories are applied at the base.
plane of symmetr
y
45 m
Fig. 8 Finite element model employed for the analysis of the case study
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The foundation is modelled with its steel reinforcement (Fig.9) and the soil with non-
linear 8-nodal continuum elements having a Von-Mises failure criterion (described sub-
sequently), while the pier column is nonlinear beam with moment–curvature (M–Cs)
relationship for actual reinforced concrete sections, and derived with the RC model of
Chang and Mander (1994). The caisson-and-piles foundation is modelled with 3D solid
elements and the concrete-damage-plasticity [CDP] model. The M-Cs relationship of this
solid Sect.(3D) is calculated with a pushover analysis, while the ultimate capacity Mu is
Fig. 9 Details of finite element modelling, including soil, foundation, and superstructure components
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calculated using RC cross-section analysis (Chang and Mander 1994). Results are shown
in Fig.10. Particular care is taken for the simulation of the failing zone, highlighted in dark
grey in Fig.8. Soil–foundation Interface behaviour allows sliding (with friction coefficient
μ = 0.5) and detachment.
Linear elastic connector elements and dashpots model the shear stiffness (Ks), the ver-
tical stiffness (Kv), and damping (C) of the isolation bearings inserted between piers (or
abutments) and deck. Appropriate gap elements, allowing adjacent nodes to be in contact
(gap closed) or separated (gap open), replicate a δc = 20 cm clearance between deck and
stoppers (shear keys), as shown in Fig. 9. This clearance is relatively small compared to
the elastic range of the bearings; hence, the assumption that the bearings remain elastic
appears reasonable (Kalfas etal. 2017), as also verified by our analyses. Calculation of the
bearing properties is undertaken using the following equations (Koh and Kelly 1988):
where A = 0.25 m2 is the plan area of the bearing (m2); t = 0.015 mm is the thickness (m)
of each elastomeric layer (Fig.9); n = 7 is the number of individual elastomeric layers;
G = 1MPa is the elastic shear modulus of the elastomeric material; Ec = 5MPa the verti-
cal stiffness of the elastomeric material; ξ = 10%t is the damping factor of the bearing; and
ω = the angular (circular) frequency in rad/s, which in this study is taken equal to the fun-
damental natural frequency of the bridge.
The deck is modeled with elastic hexahedral continuum elements having the properties
of RC (E = 30GPa, γ = 25kN/m3). We properly simulate the kinematic constraints imposed
(2)
K
h,b=
GA
tn
(3)
K
v,b=
E
c
A
tn
(4)
C
h,b=
2K
h,b
𝜉
𝜔
(5)
C
v,b=
2K
v,b
𝜉
𝜔
0
100
200
300
400
500
600
700
800
900
00.00050.001 0.0015 0.0020.00250.003
M
(x10
3
kNm)
c
(1/m)
0
20
40
60
80
100
120
00.005 0.01 0.015
P1,2,7
P3,4,5,6
M
(x10
3
kNm)
c
(1/m)
(a)(b)
Fig. 10 Reinforced-concrete cross-sectional moment–curvature relationship plots for: a the caisson of pier
5; and b columns of all seven piers of G1 bridge
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by the continuous deck. The concrete beams and slab are in accord with the structural
drawing.
Appropriate "free-field" boundaries are used at the lateral edges: dashpots are placed at
the base (
Cb=𝜌VsA
) to simulate the halfspace under the 45m of the soil (ρ = 2.1 t/m3, Vs
the shear wave velocity, taken equal to 800m/s2, and A = the effective area of each dashpot,
function of the element size). In addition, the symmetric counterpart of the model (Fig.8)
is considered, and “node-to node” kinematic constraints are applied along its lateral edges
(boundaries), forcing two nodes to have identical displacements. This aims to simulate the
response of a plane-strain lateral box subjected to in–plane vertically propagating waves.
4.2 Constitutive models
4.2.1 Soil
Apart from the narrow zone of elements to model the slip plane, the stress-strain behavior
of the soil layers (namely, III, IV, V, VI, as per Fig.4) is described by a nonlinear kinematic
hardening model obeying the Von Mises failure criterion with associative flow rule, pro-
posed by Gerolymos and Gazetas (2005) and Anastasopoulos etal. (2011) for clays under
undrained conditions. Only three parameters are required: (a) the undrained shear strength
su
; (b) the initial, small-strain soil stiffness (
E0
or
G0
); and (c) the stiffness decrease and
the hysteretic-damping increase with increasing strain (G–γ and ξ–γ curves). The model
is described in detail in (I Anastasopoulos etal. 2011), and its parameters governing non-
linear and dissipative soil are calibrated against the Vucetic and Dobry (1991) curves, as
0
0.2
0.4
0.6
0.8
1
0.0001 0.0010.010.1 110
0
0.2
0.4
0.6
0.8
1
0.0001 0.0010.010.1
11
0
0
0.2
0.4
0.6
0.8
1
0.0001 0.0010.010.1 1
Unit V
Unit IIIUnit IV
Vucec and Dobry, 1994 (PI = 0)
Vucec and Dobry, 1994 (PI = 15)
Vucec and Dobry, 1994 (PI = 30)
This Study
γ
τ
G
Fig. 11 Results of simple shear tests for the calibration of the soil constitutive parameters: comparison with
the experimental curves of Vucetic and Dobry (1991) for low plasticity clays
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shown in Fig.11 Due to lack of borehole and test data, the parameters of the stiff ophiolite
layer (Unit VI) were taken equal to those of the stiff flysch layer (Unit IV).
This constitutive law has been extensively validated previously against physical model
and numerical tests, demonstrating its effectiveness in capturing the response of surface
and slightly embedded foundations to cyclic and seismic shaking, the performance of
piles and caissons against lateral loading, and the sliding response of slopes and retaining
systems (I. Anastasopoulos etal. 2015; Garini and Gazetas 2021; Nikos Gerolymos etal.
2015; Giannakos etal. 2012; Ntritsos etal. 2015).
4.2.2 Slip surface
An elastoplastic constitutive law, with Mohr–Coulomb failure criterion and isotropic strain
softening, models the soil in the predefined sliding surface (I. Anastasopoulos etal. 2007)—
marked in dark grey in Fig.8. For this zone, the initial, pre-yielding, soil behaviour is assumed
to be linearly elastic with stiffness calculated from the secant modulus Gs as derived from soil
element testing (Soil V, in Table2). For the post-peak softening response, a reduction of the
friction angle (φ’) and dilation angle (ψ) has been calibrated against the recorded slope dis-
placements measured in-situ since 2007. Simulation of gradual degradation from c = 5 kPa,
φpeak = 33°, ψpeak = 3° in 2007 to c = 0 kPa, φres = 18°, rest = 0° in 2019 results in numerically
computed displacements that compare relatively well with the actual measurements (Fig.12).
Fig. 12 Calibration of material properties at the predefined weak zone (slip surface): a snapshot from the
analysis showing the concentration of plastic strains, and b comparison between measured and recorded
slope displacements
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Subsequently, the calculated state reached by the slope after creeping is used as the initial con-
dition to calculate the seismic response through dynamic time-history analysis (Sect.5).
4.2.3 Structural elements
The nonlinear concrete response is simulated with the concrete damage plasticity (CDP)
model (Lee and Fenves 1998; Lubliner etal. 1989), built-in the finite-element code ABAQUS,
used for both monotonic and cyclic loading (Antoniou etal. 2020; Behnam etal. 2018; Krahl
etal. 2018; Zoubek etal. 2013). The adopted model captures the effects of irreversible dam-
age associated with the two prevailing failure mechanisms of concrete: (a) tensile cracking
and (b) compressive crushing. More details of the model constitutive laws and application are
provided in SIMULIA (2014).
The parameters ψ, ϵ, σbo/σco (or fbo/fco) and Kc of the CDP model are defined as follows:
• The dilation angle of the material (ψ) was assumed equal to ψ = 15°, which is close to a
low value for concrete (it may be as high as 50°), since a less stiff response is expected
herein.
• The parameter ϵ, the “flow potential eccentricity”, defines the rate at which the hyperbolic
flow potential approaches its asymptote.
• The ratio σbo/σco of the initial biaxial compressive yield stress to the initial uniaxial com-
pressive yield stress and,
• The ratio Kc of the second stress invariant in the tensile meridian q(TM) to that of the com-
pressive meridian q (CM) at the initial yield (0.5 < Kc ≤ 1.0).
In this research, these parameters, calculated according to Alfarah etal. (2017), are listed
in Table4.
The concrete body of the caisson and periphery piles are modelled with quadrilateral con-
tinuum solid elements and their inelastic behaviour is described with CDP. The characteristic
concrete compressive strength fck is 35MPa. The tensile strength is calculated according to
EC2:
while the Young’s modulus is estimated according to Chang and Mander’s (1994):
The post-yielding of concrete is defined by the stress—strain relationship of concrete under
uniaxial compression and uniaxial tension. In addition, the relationship of the damage param-
eters
d
t and
d
c, defining the unloading–reloading behaviour of concrete, with the tensile and
compressive strain respectively, must be defined. These parameters are calibrated against the
relationships of Chang and Mander (1994); the corresponding curves are plotted in Fig.13.
ft[MPa]=0.7
[
0.3
(
fc∕1000
)2
3
]
E
=8200
(
f
c
∕1000
)3
8=
27.4 GPa
Table 4 Parameters of the
Concrete damage plasticity
(CDP) model
Kc
𝜓
(°)
𝜎
bo
∕𝜎co
𝜀
2/3 15 1.16 0.1
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Truss elements are employed to model the S500 steel reinforcement in both longitudi-
nal and transverse direction. Perfect bonding between the reinforcement and concrete is
assumed: neither sliding nor separation are allowed. The reinforcement configuration for
each pier and caisson piles is summarized in Fig.9. The caisson longitudinal steel rein-
forcement ratio ρ1 = 0.8% and the transverse t comprises Ø40 mm/15 cm. The periph-
ery piles have longitudinal reinforcement 12Ø22 steel bars, and transverse Ø30/15. Steel
mechanical behavior is described with an elastic–perfectly-plastic law with typical proper-
ties for S500 steel; yield strength fy = 500MPa, ultimate strain εu = 15%, elastic Young’s
Modulus E = 210 GPa, Poisson’s ratio ν = 0.20.
5 Response oftheG1 bridge underseismic andlandslide actions
Our analyses are based on the fundamental simplifying assumption of a synchronous seis-
mic base excitation (see a discussion on this approximation in Sextos etal. (2003)). The
analysis accounts for both land-sliding and ground shaking. The numerical analyses are
divided into the following steps:
• Initially, a static step the gravitational forces apply to the whole model.
• Next, the angle of friction (φ’) and cohesion (c) of the pre-defined slipping surface is
reduced to
• Residual values and consequently, displacement of the slope is induced (Fig.12).
0
0.2
0.4
0.6
0.8
1
00.00010.00020.00030.0004
0
0.2
0.4
0.6
0.8
1
00.002 0.0040.006 0.0080.01
0
0.5
1
1.5
2
2.5
00.00010.00020.00030.0004
0
10
20
30
40
00.002 0.0040.006 0.0080.01
fc
(MPa)
ε
c
ft
(MPa)
εt
dc
ε
c
dt
ε
t
(a) (b)
(c) (d)
Fig. 13 Concrete Damage Plasticity model: a compressive and b tensile stress–strain curves for uniaxial
loading based on Chang and Mander (1994); evolution of damage variables for c compressive and d tensile
uniaxial loading
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• As a last step, dynamic time history analyses are performed, with an initial deformation
and stress
• State from the previous step.
In other words, the calculated state reached by the slope after displacement of the
pre-defined slipping surface is used as a new initial condition to calculate the earthquake
response. Analysis is performed by performing a static step without modelling the creep-
ing. Then, dynamic analysis of the system is performed, in which slippage at the failure
surface is taking place during shaking. This is a reasonable approximation of what is
expected years after the original creep has taken place.
The eleven (11) accelerograms adopted for each of the three (3) hazard scenarios, i.e.,
weak, moderate, and strong excitation levels, are used as input to the FE analysis. The
response of the bridge to both slope deformation and seismic shaking of the H-CMP015
record (strong excitation) are presented in Fig.14. Soil displacement reaches values of
more than 3.5m, indicating full mobilization of the landslide, as seen in the contours of
Fig.14. The horizontal displacement (U1) of the caisson foundation reaches nearly 0.5m
while its rotation (θ) reaches 0.32 degrees. Evidently, mobilization of the soil deformation
and substantial horizontal displacement and rotation of the footing are realized at t = 6.5s,
when excitation is strongest. At the same time, the maximum overturning moment on the
foundation is about 70% of its ultimate capacity. Moment–curvature plots have been evalu-
ated for all the bridge piers (P1-P7). Evidently, Pier P5 suffers the most. It experiences
a ductility demand to ultimate capacity ratio of about 0.3. The other piers, unaffected by
simultaneous soil movement, are less vulnerable to earthquake excitation. The bearings of
Pier 3 displace Δu = 12cm, less than the clearance between elastomers and shear keys.
A summary of the results from all dynamic analyses for moderate and strong excitation
levels is presented in Figs.15 and 16, respectively. Results for low amplitude earthquakes
are not presented. The two figures compare for each of the 11 motions: the maximum dis-
placements at the crest of the slope; the residual deformations of the caisson foundation;
the ratios between applied moment on the caisson and its ultimate capacity; shear strains of
the bearing; curvature-ductility demand over the ductility capacity.
The study showed that P5 is the most vulnerable pier for all seismic scenarios due to
the additional stressing imposed by soil–foundation deformations. Results for the moder-
ate shaking accelerograms that correspond to the design excitation level (Fig.15) indicate
minimal damage to the bridge. Although for a few seismic cases soil displacements reach
significant values (of the order of over 1m), foundation deformations remain limited, and
the applied moment on the caisson footing for all cases is well below its ultimate capacity;
this is indicative of elastic foundation response: thereby no damage is likely of concrete
and steel reinforcement. Ductility demand for pier 5 remains below 10% of its ductility
capacity.
The corresponding results for the strong excitations are presented in Fig.16. Now, crest
deformation reaches values of more than 3.5m (with the H-CMP015 excitation, illustrated
in Fig.14). Ductility demand on Pier 5 is higher but still within acceptable levels, while the
moment applied to the foundation is substantial, indicating concrete cracking. The above
results show the potential of severe damage to some bridge components if a relatively
strong earthquake excitation is accompanied by full landslide mobilization.
Overall, the "as-built" bridge indicates that the foundation remains within its elastic
structural regime, with minimal damage even for the worst-case scenarios, suggesting that
a more robust foundation would be seismically unnecessary. In addition, the deck bear-
ings remain within their elastic limit, and displace less than the clearance limit (of 20cm).
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As a consequence, the shear keys are not engaged. Notice that excitations H-CMP015 and
SSE330 are the most pernicious as they contain a large number of excitation cycles.
However, on the contrary, the simultaneous mobilisation of the landslide leads to
permanent footing rotation and increased damage of the pier and superstructure. Hence
Fig. 14 Response of the G1 bridge under a “strong” seismic excitation (H-CMP015 record) interacting with
slope sliding deformation: a worst-case bearing differential displacement; b moment–curvature relationship
for P3 and P5; c horizontal displacement and d rotation at the top of the foundation
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Fig. 15 Comparison of the response of foundation and bridge for the 11 moderate seismic excitations
(10% in 50years): a Slope crest maximum horizontal displacement; b Caisson foundation top horizontal
displacement; c Caisson top rotation; d Maximum overturning moment on the caisson over its ultimate
moment capacity; e Maximum shear deformation of the elastometallic-bearing; f Maximum curvature duc-
tility demand over ultimate ductility for the column of pier 5
Fig. 16 Comparison of the response of foundation and bridge for the 11 strong seismic excitations (2% in
50years): a Slope crest maximum horizontal displacement; b Caisson foundation top horizontal displace-
ment; c Caisson top rotation; d Maximum overturning moment on the caisson over its ultimate moment
capacity; e Maximum shear deformation of the elastometallic-bearing; f Maximum curvature ductility
demand over ultimate ductility for the column of pier 5
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enhancing the robustness of the foundation would be necessary. To this end, an alternative
scheme to improve the performance of the system by installing “nailing” pile rows uphill
of the foundation (in place of the questionable peripheral pile ring). If effective, apart from
their stabilizing effect, such piles would also act as sacrificial members, protecting the cais-
son foundation in case of extreme shaking. The response of such an alternative is analysed
below.
6 Response afterinstallation ofmitigating piles
6.1 The need formitigation
To further elucidate the negative role of slope instability on the performance of the system,
we decouple the effect of land-sliding from that of the terrain topography. To this end, we
ignore the presence of slippage interface, i.e., the upper clayey–silty-sand layer (Unit V)
lies directly onto the flysch–and–ophiolite layer (Unit III, Unit IV and Unit VI). We con-
sider the initial ("as-built") scenario only. Comparison of foundation response of the origi-
nal analysis and this new no-landsliding analysis is presented in Fig.17.
The comparison reveals the significance of slope instability on the response of the
bridge. It appears that seismic shaking alone—i.e., kinematic effects, inertial loading, and
even “topographic amplification” (aggravation of amplification from the non-horizontal
(a) (b)
(c) (d)
Fig. 17 a Contours of horizontal soil displacement due to GYN000 excitation where the predefined slip
surface is neglected, and b of the original case where slippage takes place; c horizontal displacement and d
rotation of the foundation top for the above two scenarios
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ground surface, see Assimaki and Gazetas (2004))—lead to only about one-half of the
total deformation of the system. Hence, stabilizing the upper layer by nailing the slippage
zone is justified to reduce soil movement. This would improve the overall response, and
perhaps could have allowed a reduction of foundation size, if it had been adopted before
construction.
6.2 Alternative mitigation measures
We propose "nailing" the sliding surface with rows of piles, while at the same time remov-
ing (as unnecessary) the periphery piles of the caisson-piled foundation of P5. The effec-
tive total diameter of the footing reduces from 9 to 7m.
We explore the effectiveness of two different pile position scenarios, each with two
additional double-pile frames: (a) the two pile frames are installed at 47m and 74m dis-
tance from the center of the caisson, and (b) the two pile frames at 17m and 47m distance
from the center of the caisson (Fig.19). Each frame consists of 1.2m-diameter piles, heav-
ily reinforced longitudinally (ratio of ρ1 = 2% of the pile cross-section). The piles, spaced 3
and 4 diameters apart as shown in Fig.19, have a length of about 35m, ensuring adequate
penetration into the stable layer below the slip surface (approximately 10m depth), and
thereby allowing full flexural mobilization of the pile-group capacity as well as activation
of the ultimate reaction to the stabilising ground. They are modelled with non-linear beam
elements the nodes of which are connected to the surrounding toroidal solid element nodes.
The heads of the piles are connected rigidly, simulating the effect of an RC pile cap. The
piles are installed in a staggered configuration that minimizes the shadow effect created
from the front row of piles (Poulos and Davis 1980), consequently offering a substantially
increased total reaction force to the system (Kourkoulis etal. 2011).
Figure 18 illustrates the slope deformation contours used as a guide for selecting the
proper mitigation measures, and Fig.19 the proposed pile-row positions for the two alter-
native configurations. Configuration A is placed in the area of the maximum soil defor-
mations. The location of Configuration B is meant to intercept the soil movement at the
middle the failing slope and the toe of the landslide, close to the caisson. The effective-
ness of each scheme is evaluated in Figs.20 and 21 for the worst-case scenario (GYN000
excitation).
Fig. 18 Slope deformation contours following the GYN000 strong earthquake excitation. Rehabilitation
measures have been based on the soil displacement distribution along the slope
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The results unequivocally show that the performance of the system with both mitigation
measures, A and B, has improved despite the reduced foundation size: ground displace-
ments have been reduced substantially and the overall performance of superstructure and
foundation is enhanced. Location B is superior for the caisson, location A for the slope.
With configuration A, soil movement at the crest substantially decreases to only 0.6m,
from 1.6m of the "as-built" case. Caisson deformation, however, remains at similar levels
with the initial case; hence, no significant improvement of the bridge behaviour is achieved.
This is because land-sliding is restrained at the upper part of the slope, but localized soil
failure at the toe is barely reduced. Hence, the caisson behaves in a similar way with the
unimproved case.
With Configuration B, displacement of the crest is 0.8m, larger than in A, but the foun-
dation behaviour is improved in terms of both rotation and horizontal displacement. This
stems from the interception of landslide mobilisation just before the caisson.
With both pile locations (A and B) the distress of the pier (P5) and the moment onto
the caisson decrease by approximately 20% with respect to the initial system (Fig.21).
4D
3D3D
3D
(b)
(c)
(a)
Fig. 19 a Plan view of staggered pile-row set-up; b Positioning of pile-row Configuration A; b Positioning
of pile-row Configuration B
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The piles acting as a remediation feature reach their ultimate capacity during the strong
event, hence providing the maximum possible reaction force. Thus, these piled-frames
serve as sacrificial members that protect the structure from seismic and landslide-
induced actions. Pile systems that reduce the slope movement at the toe (B), while
allowing full mobilization of slippage near the crest (type B), are the most beneficial.
The installation of periphery pile is not helpful.
-0.4
-0.2
0.0
0.2
0.4
0510 15 20 25 30
a
time
GYN000
δ
soil
(m)
U
1
(m)
δ
soil
(m)
toe
θ(deg.)
crest
-0.2
-0.18
-0.16
-0.14
-0.12
-0.1
-0.08
-0.06
-0.04
-0.02
0
0.02
0510 15 20 25 30
0.00
0.05
0.10
0.15
0.20
0.25
0.30
0510 15 20 25 30
-1.6
-1.4
-1.2
-1
-0.8
-0.6
-0.4
-0.2
0
0510 15 20 25 30
-1.6
-1.4
-1.2
-1
-0.8
-0.6
-0.4
-0.2
0
0510 15 20 25 30
(a) (b)
(c) (d)
δ
soil
(m)
δ
soil
(m)
time (s)
cresttoe
U
1
(m)
θ
(deg.)
Current State
Conf. A - Migaon
Conf. B - Migaon
Fig. 20 Comparison of the two interventions on the response of the slope–caisson system to GYN000 exci-
tation: a Horizontal displacement of the crest; b Horizontal displacement of the toe; c Horizontal displace-
ment of the caisson top; d Rotation of the caisson
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7 Conclusions
The behaviour of the G1 bridge under multi-hazard loading (earthquake shaking and land-
sliding actions) has been evaluated numerically. The results—including slope residual
movement, foundation displacements, and superstructure response—show that ground
motions of an exceedance probability of 2% in 50years would trigger land-sliding defor-
mations, thereby generating significant residual deformation of the foundation and struc-
tural damage to the piers. Decoupling the effects of sliding and vibration reveals that the
effect of the former is critically important, and its mitigation should be a priority.
An alternative bridge foundation strategy is developed to reduce the land-sliding, rather
than to make the foundation more robust. The piles along the periphery of the caisson are
removed (reducing the total foundation diameter from 9 to 7m), and staggered-pile-frames
are installed to minimise the potential slope movement. Two possible locations of such
pile-frames are studied, one emphasizing the “arrest” of the crest displacement (location
A), and the other the ground displacements in the vicinity of the foundation (location B).
Both interventions are successful. Unsurprisingly, location A leads to smaller slope dis-
placement at the crest, but it is location B that offers the largest benefit to the foundation
and the structure.
Acknowledgements This research has been financed by the European Commission through the Horizon
2020 program "PANOPTIS—Development of a decision support system for increasing the resilience of
transportation infrastructure based on the combined use of terrestrial and airborne sensors and advanced
modelling tools", Grant Agreement number 769129. The first author acknowledges the financial support
provided by the research and innovation programme under the Marie Sklodowska-Curie grant (grant agree-
ment No INSPIRE-813424, "INSPIRE—Innovative Ground Interface Concepts for Structure Protection").
The authors are grateful to Dr. Panagiotis Panetsos of Egnatia Motorway for providing all the necessary data
for the case study investigation. The authors would also like to thank the reviewers for the truly valuable
constructive comments that helped improve the quality of this paper.
Funding Open access funding provided by HEAL-Link Greece. This research has been financed by the
European Commission through the Horizon 2020 program "PANOPTIS—Development of a decision sup-
port system for increasing the resilience of transportation infrastructure based on the combined use of
Fig. 21 Comparison of the two interventions on the response of the slope–caisson system to GYN000 exci-
tation: a Moment–curvature relationship for the column of P5; b Time history of maximum moment in the
caisson normalized by its moment capacity
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Bulletin of Earthquake Engineering
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terrestrial and airborne sensors and advanced modelling tools", Grant Agreement number 769129. The first
author also acknowledges the financial support provided by the research and innovation programme under
the Marie Skłodowska-Curie grant (grant agreement No INSPIRE-813424, "INSPIRE—Innovative Ground
Interface Concepts for Structure Protection").
Declarations
Conflict of interest The authors declare that they have no conflict of interest.
Open Access This article is licensed under a Creative Commons Attribution 4.0 International License,
which permits use, sharing, adaptation, distribution and reproduction in any medium or format, as long
as you give appropriate credit to the original author(s) and the source, provide a link to the Creative Com-
mons licence, and indicate if changes were made. The images or other third party material in this article
are included in the article’s Creative Commons licence, unless indicated otherwise in a credit line to the
material. If material is not included in the article’s Creative Commons licence and your intended use is not
permitted by statutory regulation or exceeds the permitted use, you will need to obtain permission directly
from the copyright holder. To view a copy of this licence, visit http:// creat iveco mmons. org/ licen ses/ by/4. 0/.
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